I remember when Module 6 came out many of the engineers in the office were surprised to see such a low load factor for Q for the static case. Whilst I am not a Geotech, I understood that the static case in the guidance was intended to represent a "drained soil case" where the footing was exerting pressure on the founding soil for a long period of time. In this context, the use of a live load coefficient of 0.4 was in line with the use of a long term live load factor, ψl. So it made sense to me in that way.

I guess I'm just thinking that the original guidance would not have contained such an obvious flaw without some kind of rationale to justify it. Wondering what others think about this one. 

So far I know it's a proprietary product cast into a wall panel to connect hollowcore floors and/or topping slabs.

Attached is the sort of detail I have, and that's about all I can find on google too.

The building I'm assessing already has seating retrofitted, so primarily I need to know if these things provide any additional capacity to transfer diaphragm loads into the wall.

There is a paragraph in clause 5.4.1 of the wind code "Under-eaves pressure shall be taken as equal to those applied to the adjacent wall surface below the surface under consideration"

My question is, should this be added to the pressure acting on the top of the roof (the "over-eave")?

I've got a fairly typical warehouse with a low-pitch roof, which extends out 2m over the wall where the roller doors are. I'm using +0.7 as the under-eaves shape factor (Table 5.2a) and -0.9 as the over-eaves shape factor (Table 5.3a) for a net shape of 1.6 upwards.

I'm comparing it to the canopy shape factors (Table B9) which have a max of 1.5 upwards, and wondering if I'm being too conservative?

First problem I found is you can't edit the front slope, which means I can't use the software for my current project (a wall halfway up a slope). As I understand it, having a front slope drastically reduces your passive resistance and affects the solution. 

Second item is more of a "it would be nice to have this". I see that there is only 2 soil layers (backfill vs founding soil) and the change is assumed to occur at the base of the cut face. It would be nice to have the analysis capable of 2+ soil layers at varying height.

I have a bit of a curly question that i thought might be worth putting out here to see if anyone else has come across the same consideration before.

I have an existing 1950's in-situ concrete 2 storey building with part height basement that I am tasked with doing the structural upgrade design for. The building is currently at approx. 35%NBS IL2 and the client wants to upgrade to 100%NBS.

The plan for the upgrade is to demolish the existing building down to ground floor level, keeping one of the external walls as it is a party wall. The existing deep foundation piers are to remain. As the building has an extensive foundation structure, i would estimate over half of the buildings original mass is to remain. The upgrade involves a new 2 storey lightweight timber/steel super structure, to be supported off the existing ground floor and foundation structure. 

My question is: What capacity parameters should be used for the existing structure? Are probable capacities okay to use (like in assessment) or should design capacities be used (as for new design). As the result is quite significant, especially for the soil interaction which generally goes from a design reduction factor of 0.5 to a probable which is set at 1.0 (as per section C4.5.2 of MBIE guidelines 2017).

The new design component above ground floor, will be done with design capacities, that is clear to me. 

Interest to hear if anyone else has asked themselves this question before and where the line sits between strengthening and new design. Appreciate any feedback. 

This query is specifically with reference to Appendix C5D, of the C5 engineering assessment guidelines for concrete buildings.

Specifically with regards to the modified Hrenikoff model analysis methodology documented therein including recommendation for section sizes to obtain an elastic solution of forces in the diaphragm elements.

It is my understanding that this methodology / Appendix has been published in the assessment guidelines prior to Amendment 3 (August 2017) of the NZS 3101:2006, at a time when reinforcement with a lower ductility class than E was allowed to be used for some reinforcement.

This methodology / Appendix has also been in place historically prior to the recommendations in the assessment guidelines to ignore cold-drawn steel mesh (often, the major typical reinforcement content in some historical buildings).

While the methodology detailed in Appendix C5D references a paper by Holmes (2015) it is listed as unpublished in the references list. (I have not found this paper in my own searches).

I am making an educated interpretation or guess that the adaption of the Hrenikoff analysis methodology by Holmes had the purpose of estimating forces at relatively small yields in order to make use of historical reinforcement such as cold-drawn welded mesh for example. Thus it is technically not a strut-&-tie analysis.

The issue I am finding is that predominantly engineers in the industry including peer reviewers and territorial authorities have adopted the interpretation that the elastic solution only to the modified Hrenikoff model is the definitive methodology for all Strut-&-Tie models (as per Appendix A of NZS3101) only including new designs with ductility class E steel.

At the recent, 2024, SESOC / HERA seminar titled “Seismic Design of Diaphragms in Steel Frame Structures” this interpretation was essentially reinforced as this method was described as equivalent to the Strut-&-Tie method. I believe this is a little misleading as it really is potentially only one of an infinite number of exact solutions since Strut-&-Tie is a plastic analysis methodology based on the lower bound theorem. My understanding is that well detailed and carried out, with appropriate materials (e.g. ductility class E reinforcement) a Strut-&-Tie solution as per Appendix A of NZS3101 already should provide a conservative and safe design or assessment solution.

It also appears to me that in some situations the limitations of the elastic solution in the modified Hrenikoff model may be both unduly conservative and unrealistic given that in a squat element, “D” region, such as a floor plate diaphragm or squat reinforced concrete wall one would expect some in-elastic behaviour at ULS and MCE scenarios and this is OK. In the context of diaphragm design and/or analysis for new design or assessment of existing floor plates we are already typically using design loads for the diaphragm factored up by the building over-strength (pESA) method, which is ensuring adequacy of design for the diaphragm floor plate.

I understand I have missed the opportunity for public consultation for the proposed revision of non-epb seismic assessment guidelines.

However, I am wondering whether it may be taken into considerations that in future revisions of the seismic assessment guidelines, Appendix C5D may add some background from the original authors on the original intention of the methodology e.g. was it intended for assessment purposes of diaphragms with lower strain capacity reinforcement, very low target crack widths and/or with addition of retrofit solutions with limited strain capacity?

Also some guidance on what is the relationship of the modified Hrenikoff analysis method to Strut-&-Tie method in Appendix A of NZS3101 would be helpful (since it is now being commonly adopted as the definitive interpretation of strut-&-tie in the NZ structural engineering industry).

Who would I contact with regards to the non-epb seismic assessment guideline revisions, with regards to this?